Overturning of Masonry Shear Walls

[bvbuf]

Problem: Three story non load bearing masonry shear walls. Control Joints at 25' max per Masonry Design Guide. Thus the maximum shear wall length is 25'(and maximum resisting moment arm is 12.5'). How can I qualify these walls in overturning without using a very large wall footing or pouring a large chunk of concrete at the ends of the shear walls for the dead load resisting moment.
Also, per IBC2000 no mention is made of using a 1.5 factor of safety for overturning of masonry shear walls thus I am using Load combination formula 16-12 0.6D+0.7E.

Any help is greatly appreciated.

bvbuf

 

[JAE]

For non-load bearing shearwalls, your only option is to anchor the wall with heavy foundations or some kind of deep foundation tie-down such as a properly reinforced drilled pier with bell or other tension-capable pile.

Your tension "arm" is not 12.5 feet but the distance from the compression "block" on one end of the wall and the tension reinforcing centroid at the other....similar to a reinforced concrete beam.

Grouting the walls full adds to the weight of the wall and helps but may not be enough for the total overturning.

I don't have my IBC copy with me now so can't respond to the 1.5 factor....even if it is not in the code, it is good practice to have this 1.5 factor working for you.

 

[lsmfse]

I agree with JAE- It is not uncommon to use drilled piers(caissons)or heavy ftgs. Also, you might consider 4' or 5'or whatever foot wide footings that extend beyond the wall. This not only helps with the RM but helps with soil brg as well.

 

[bvbuf]

Thanks JAE
Thanks Ismfse

I really appreciate your time.

I agree it is good practice to use the 1.5 FS for overturning but the project is in Georgia and the 2000IBC EQ provisions (we believe) are unnecessarily high for this region. Thus until we get updated loading (and we hear it is coming soon) I don't want to go beyond code mandated minimums.

Some other ideas have come up..
One suggested that since some shearwalls are immediately adjacent, I can take advantage of the fact that during a seismic event when wall A has a upward force on the footing, wall B (coplaner and immediately adjacent) has an downward force on the same footing. Thus they would cancel each other. (The walls are stitched together at the floor level and the roof level thus they should move together.)

A second idea would involve extending the rebar in the bond beams across the expansion joint. Thus I could use the dead load of the adjacent panel to help with uplift at the expansion joint.

Any thoughts?

 

[JAE]

I agree that the adjacent downward force counteracts the upward force in the other wall. However, this load path goes through your footing at the joint in shear through the footing. Wall A is pushing down on the footing and right next to it Wall B is pulling up....high shear no doubt.

A bond beam would probably not be able to develop the high shear that would occur.

Higher seismic forces are developing all over due to more research and thought being invested in the statistical EQ data. This requires a lot more of our efforts in design - with probably little extra compesation fee-wise.

 

[ronster]

In response to the overturning SF, you are only allowed to use .6xDL against full wind load or .7x seismic loading in allowable strength design (ASD). The overturning SF is 1/.6 or 1.67 which comes from your reduced dead load. Do not had a safety factor on top of this. Earthquake loads are already factored, you use 1.0E in LRFD and .7E in ASD. Does anyone know why earthquake loads are factored but wind loads are not in the IBC? Is it part of the bigger plot to try to confuse everyone with the lateral load analysis so that everyone gets to work more for the samer pay? Or is that the people who write these codes spend a little to much time in schools abstracting trying to justify their work with complicated formuli and not enough time in reality? Pardon my rambling.